huangyhg

help with beam design
i am trying to design the steel beam shown in the attached sketch. the beam span is 34'-0" and is loaded with 300 pounds per foot from the floor (one side only) and with 700 pounds per foot on the top flange from the roof framing (snow loads included). because the top flange is unbraced i am coming up with a w30x99 (fy 50)!! i don't think i can get the floor diaphragm to brace the top flange because the connection would have to act in tension or compression depending on which way the beam wants to rotate. any suggestions?
you need to provide some full height stiffeners between the top and bottom flanges.
alternatively you could use a lower beam with the top flange at floor height.
csd72-
are you implying that full height stiffeners will act as a brace?
you could run a plate from flange to flange parallel to the web (on the side that you do not have the joists framing into) and stitch weld the two together. this should give you the rotational stiffness to resist the ltb. otherwise, consider it unbraced.
that's not an unreasonable size you came up with. i would go with unbraced and keep it simple.
does the 2x12 on the right side recieve a floor?
yes, the 2x12's are at 16"oc and carry 30psf live and about 15psf dead and they span about 13'-0".
sorry steve, i re-read your post and you did say the floor was only on one side. i'm not sure why you would need the 2x12 on the right, i thought maybe something was coming in on that side, too.
i agree with ucfse and would go with the unbraced length and keep it simple. but i did do a quick calc for this (asd) with 1.1 k/ft and unbraced length = 34'. with w18x86 i get fb = 11.35 ksi, fb = 13.67, and deflection ~ 3/4" (l/544). are you trying to get less deflection? concern for torsion?
your detail currently shows a beam only about 18" deep, but you need 30"? is this detail the architects?
you should be able to install a network of full height vertical and horizontal stiffeners on one side and weld straps periodically from the stiffeners to be nailed off to the plywood floor diaphragm. depending on how often you install the vertical stiffeners and straps, that should establish the laterally unsupported length. with a 34'span and a 16" joist spacing, a max of 5'-4" should work.
i wonder how you will address (finishing)the flange width at the intersection to the top wood bearing wall of the w30 you mention which is 10.5" as opposed to a 5.5" wall with sheetrock/plywood? i think that you are going to have to comeupwith a beam with a smaller flange width for the architect to buyoff on it.
upon looking at the numbers you presented, the load is 1 kip per foot with a 34 foot span, generating a simple span moment of less than 145 kip feet. if you are braced at midspan, a w18x50 should work with a 7.5" flange, but have a 1.3" deflection (l/316)(aisc 13th edition table 3-10, page 3-125. am i missing something here?
mike mccann
mccann engineering
i apologize . . . the floor is on the left only. the 18x86 might work but the 18x50 will not since the mispan cannot be braced. adding the selfweight of the beam produces a bending moment of 157 ft kips. even if the midspan could be braced, the bending stress of the w18x50 would be 20.47ksi which exceeds the allowable of 15.37ksi.
or put a steel channel on top of the beam (like a bridge crane beam) to add top chord rt value and possibly get a smaller net section (in terms of weight). the cost of welding on a continuous channel might be smaller than the cost of a bunch of vertical stiffners.
and i'd worry that the vertical stiffners might someday get "mis-perceived" and removed, altered or otherwise made ineffective.
then we'd have "fall-down-go-boom" mode of failure.

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